Category HIGHWAY ENGINEERING HANDBOOK

Adequate drainage of the deck is important for safe operation during rainstorms, to prevent accumulation of rainwater or snowmelt that could freeze and cause skidding, and to prolong the life of the deck by removing standing water, which would otherwise contribute the water element necessary for corrosion. Transverse drainage of the deck should be provided by cross-slope whereas zero gradients and sag curves on bridges should be avoided.

Design of the drainage system is frequently one of the last items listed in the scope of ser­vices for a bridge design project. This implies a secondary importance, but that is certainly not the case. On the contrary, the South Carolina DOT requires that scuppers of adequate size and spacing for high-intensity rainfall be shown in the preliminary bridge plan view. The DOT reviewers check the scupper design as part of the preliminary design review. This is the stage when deck drains should be designed. At this stage changes as major as span lay­out revision can be made with relative ease to ensure adequate drainage, including provision for longitudinal conductors with acceptable slopes, and downspouts, if necessary.

Scupper Size and Shape. Bridge scuppers are unlike roadway drains, which can have large grates through which the storm water flows and drops into a large sump and then is conducted away by transverse or longitudinal drain pipes. Bridge scupper outlet pipes generally must be small circular, square, or rectangular pipes, and so any enlarge­ment of the scupper surface opening must be limited to prevent debris from being trapped, clogging the scupper and making it ineffective. A minimum surface opening larger than the diameter of a beverage can is desirable and should be maintained to the outlet end. Some states require the use of bicycle-safe grates on bridges where bicycle traffic is allowed. Bicycle-safe grates have smaller individual openings that would pre­vent catching of the tire but should provide the same overall opening area required by the drainage design. An open 4-in – (100-mm) or 6-in-diameter (150-mm) pipe meets this requirement and is economical compared with scuppers having large fabricated boxes with crossbars, so that more of them can be provided at the same cost. Where the dis­charge can be corrosive, such as where deicing salts are used, the pipe should extend below the bottom of an adjacent beam to prevent the discharge from being blown onto the beam. This is especially important on weathering steel bridges, and is also important on painted steel and concrete bridges. In areas that do not experience freezing, a formed opening in the deck may be acceptable, or the pipe need not extend below the beam. In this case one can expect a stain to develop on the beam onto which the discharge is blown. Some states prohibit drainage directly onto unpaved embankments or natural ground where erosion could undermine structural elements, or onto any traveled way, either vehicular or pedestrian.

Deck Drainage Criteria. Deck drainage design is based on preventing storm water from spreading in the gutter more than an acceptable amount during a rainfall of a given intensity. For example, the maximum permissible spread may be the width of the shoulder, or the spread may be permitted to extend across half of the outside lane of a multilane directional roadway. One source of such drainage design criteria is the U. S. Department of Transportation document “Drainage of Highway Pavements,” Hydraulic Engineering Circular 12 (HEC 12).

Scupper Design Procedure. After the size of the scupper opening is established, the spacing may be determined. It’s a common practice to locate the scuppers just ahead of the bridge joints to collect most of the runoff water therefore reducing the possibility of rainwater leaking through damaged bridge joints and causing deterioration in super­structure and substructure. In bridges with longer spans, rather than try to directly determine the spacing, it may be easier to select a trial spacing and then check the ade­quacy of that spacing by using a hydraulic analysis method acceptable to the client. The use of computer programs greatly expedites this task.

Collection of Runoff. In areas such as reservoirs and sensitive wetlands, it may be necessary to provide a collector system and temporarily detain the first half inch or inch of rainfall from each storm. It is assumed that the initial rainwater that falls at the beginning of a rainstorm will contain most of the roadway pollutants that can be carried away with the runoff. The objective is to prevent these pollutants, especially petro­chemicals (crankcase drippings, fuel spills, etc.), from polluting the area under the bridge. The collected storm water is treated and then discharged. In some cases this protection is extended to the roadway as well.

Repainting of bridges over highways or railroads may necessitate use of protective covers, or require traffic lanes to be diverted or work interrupted during passage of trains, while existing paint is removed and new paint applied. These factors all favor use of a bridge material or protection system that does not require maintenance reapplication of a coating. (See Art. 4.13.)

Removal of Existing Paint. Complete removal of the existing paint on a bridge that is to be recoated with a paint system that requires it can be extremely expensive, particularly if the existing paint contains lead. Lead-based paints were used extensively in the past because they provided good protection. Because of the health issue involved, portions of bridges where lead-based paints are being removed are often required to be completely enclosed, and the paint particles contained and properly disposed of, often at great cost. Severe monetary penalties can be imposed if violations occur. (See Art. 1.4.)

Repainting. Maintenance paint should be applied with the same care as paint on new bridges, but it must be applied in a more difficult environment. Painting must be done under acceptable atmospheric and environmental conditions, particularly in regard to temperature, humidity, wind, and absence of dirt. Overspray onto vehicles and other objects must be prevented. Some painting contractors are consciously careless about this, preferring to take their chances and let their insurance company pay claims, rather than taking necessary precautions. This results in bad public relations between travelers and the owner.

Because of the cost associated with complete removal of existing paint, paint systems that do not require complete removal, but only removal of loose paint and minimal preparation of sound paint and exposed steel, are much desired. Some such systems are on the market. While longevity can be projected by accelerated testing, only real­time exposure will truly prove their worth.

Inspection. Thorough inspection during repainting contracts is essential to satisfactory performance. Painting contractors often work during off-hours, and so the owner’s inspectors should be prepared to work those same hours. Inspectors should follow closely behind the painters. On high bridges, the use of inspection devices such as high lifts, Reach-Alls, Snoopers, or cherry pickers, which permit inspectors to reach areas otherwise not readily accessible, will keep the painters on their toes just through awareness of their availability to the inspector.

Shoulders were not always provided on bridges in the past. This in itself can be a reason for widening an existing bridge. More frequently, widening is necessitated by the addition of lanes to the highway, at which time a full shoulder can be provided.

The design and preparation of plans for bridge widening usually require all the same elements as the preparation of the original plans for the structure, plus details and notes for partial removal of the existing bridge, rebar splice details, and notes on sequence of construction and maintenance of traffic. Therefore, it is a mistake to think of such a design project as “just a widening job” when estimating the hours required to design and prepare plans, or when reviewing such estimates for agency approval.

Bridges are generally widened in kind—that is, steel-beam bridges are widened with addi­tional steel beams, prestressed-concrete beam bridges are widened using prestressed-concrete beams, etc. However, beam types different from the original have been used successfully in some widening projects.

100 125 150 175 200 225 250 275 300 325 350 375 400 425 450 475 500 Span length, feet

—■— Superstructure cost —•— Substructure cost —►— Total cost

FIGURE 4.11 Example of cost study for optimum span length. Conversions: 1 ft = 0.305 m, 1 sq. ft = 0.0929 m2.

It is sometimes possible to increase the design load capacity of a bridge when widening. If the bridge is steel and was originally designed and constructed noncom – positely, it may be feasible to weld shear studs onto the top flange of the beam or girder if the deck slab must also be replaced. If the deck slab is good, another available tech­nique is to carefully core holes in the slab over the beams, weld shear studs onto the beam flange, and fill the hole with high-strength concrete, thereby making the beam composite. One should always be cognizant of the effect of retrofits on fatigue life, just as one is conscious of the fatigue effect of structural details on new construction.

Another means of increasing load capacity is to space the existing beams closer together. Before this is done, a study should compare the cost of renovation with the cost of replacement with a new superstructure.

Posttensioning of members can also be used to increase the load capacity of existing bridges, or to correct deficiencies in the original design. External posttensioning of prestressed or posttensioned girders has been necessary on some bridges where the design did not adequately anticipate the magnitude of time-dependent deflections that occurred. In one case, a utility bridge developed a sag that trapped rainwater, further increasing the deflection. It was corrected by external posttensioning. Members of truss bridges can be posttensioned to increase the load capacity of the truss. A com­puter program is available from the BEST Center, University of Maryland, which allows analysis including the effect of posttensioning cables.

When determining the load capacity of an existing bridge, one should refer to the original plans, if available. These plans will generally state the design specifications used and the type and required strength of materials. For steel bridges for which plans are not available but the year of construction is known, the type of steel and the allow­able stress may be obtained by reference to the AASHTO Manual for Maintenance Inspection of Bridges. If the bridge is a large or significant one, the type and strength of the steel should be determined by chemical and physical analysis performed on a coupon taken from the bridge. The chemistry, particularly the carbon equivalent, will be important if welding is proposed on the existing steel.

When evaluating the strength of an existing bridge for widening or rehabilitation, refer­ence should be made to AASHTO publications dealing with evaluation and rating for strength and fatigue. Two such publications are the AASHTO Guide Specifications for Fatigue Evaluation of Existing Steel Bridges and the AASHTO Guide Specifications for Strength Evaluation of Existing Steel and Concrete Bridges. Manual for Condition Evaluation of Bridges and Guide Manual for Condition Evaluation and Load and Resistance Factor Rating (LRFR) of Highway Bridges are also available through AASHTO.

Evaluation of an existing prestressed-concrete girder bridge for which plans are not available may be more difficult. The type of girder, whether a standard AASHTO shape or a state standard, may be determined by measurement. The number and size of strands may be apparent at an exposed end, but whether any strands are deflected or debonded, whether the strands are stress-relieved or low-relaxation, and what their strength is are not easily determined. Full-scale load-deflection testing may provide some answers, but is very expensive. This illustrates the importance of maintaining and safeguarding the original plans and as-built drawings, and having complete design data on those plans.

Where the spans are not controlled by features crossed—such as roads, railroads, streams, or existing buildings—and there is freedom to locate piers, the lengths of spans will be controlled by aesthetic, economic, and structural requirements. Generally, from an aesthetic standpoint, spans should have a length at least 3 or 4 times the pier height.

The profile of the site crossed will influence the span proportions. On the uphill end of a crossed hillside, the end spans will be shorter than at the bottom of the valley. The type of bridge will also affect the selection of span ratios, from both aesthetic and structural standpoints. Where spans are continuous, the end span should not be made too short, because uplift may occur under live load, and loss of positive reaction at the abutment will occur sooner if the abutment settles.

The most economical bridge will generally not be either the one with the most eco­nomical superstructure or the one with the most economical substructure, but the one with the least combined cost. That determination is made by performing a cost study wherein a number of different span lengths are investigated, along with the cost of their substructures. To be meaningful, the superstructure and substructure designs should be fairly detailed. The superstructure and substructure costs are then plotted. The optimum span length will be at the low point of the combined cost curve. A typical cost study curve is shown in Fig. 4.11.

The steel box girder bridge is depicted in Fig. 4.10. The steel elements are fabricated and erected as “tubs,” and the composite concrete deck is placed in the field. This con­figuration has some advantages over plate girder construction. Visually, it is “cleaner,” and it does not provide surfaces for birds to perch. For high-visibility bridges, such as urban interchange bridges where motorists are constantly passing beneath the bridge, the enhanced appearance may be a deciding factor. Also, the cleaner surface areas tend to improve durability and reduce repainting costs. The bridge is torsionally stiff—especially

beneficial for horizontally curved bridges. These advantages come at a price, however, because box girder bridges are generally more expensive than plate girder bridges. Sometimes the extra cost is knowingly borne for the aesthetic advantage.

In regard to efficiency in the number of lines of girders in bridges consisting of multiple girders connected by cross frames, cursory cost comparisons almost always conclude that the widest spacing of girders is the most economical. Savings result not only from the reduced number of main members but also from the reduced number of secondary elements (shear connectors, cross frames, stiffeners, and bearings). However, other costs must be considered. Wide girder spacing will generally be accompanied by a wide slab overhang over the outside girders, for a balance of load on interior and exterior girders. This may necessitate extra reinforcing steel in the top of the deck slab beyond the amount required for the slab span between girders, and may require a thicker slab. Use of three lines rather than four or more puts the bridge closer to a nonredundant condi­tion. In some cases, a greater number of girders than the optimum for minimum material cost may be necessary or desirable to permit the bridge to be built in stages, or to have the deck replaced while maintaining traffic on a portion of the deck width. In general, beam and girder spacings of up to 10.0 or 12.5 ft (3 or 4 m) should be investigated for typical bridges. For economy, the size of interior and exterior stringers should be the same.

The concrete deck for steel beam and girder bridges may be designed and constructed on the basis of either composite or noncomposite behavior. With composite construction, the effective area of the slab can be calculated and used in determining the moment resistance of the section in positive moment regions. In negative moment regions, ten­sile stresses can be resisted by the reinforcing steel. The required number of shear connectors must be calculated and furnished. These are generally headed studs that are welded to the top flange (Fig. 4.9). Overall economy depends upon the cost of the installed shear connectors and the reduction in steel weight that can be obtained. However, composite construction is frequently the economical choice.

4.16.1 Economical Design of Steel Plate Girder Bridge

Suggestions for maximum economy of steel girder bridges may be summarized as follows[5]:

1. Load-and-resistance factor design (LRFD) is the preferred design procedure. Load-factor design (LFD) yields more economical girder designs than does allowable – stress design (ASD).

2. Properly designed for their environment, unpainted weathering-steel bridges are more economical in the long run than those requiring painting. Consider the fol­lowing grades of weathering steels: ASTM A709 grade 50W, 70W, HPS70W, or 100W. Grade 50W is the most often used.

3. The most economical painted design is that for hybrid girders, using 36-kip/in2 (248 MPa) and 50-kip/in2 (345 MPa) steels. Painted homogenous girders of 50-kip/in2 (345 MPa) steel are a close second. The most economical design with high-performance steel (HPS) will also be hybrid, utilizing grade 50W steel for all stiffeners, diaphragm members, and web and flanges, where grade 70W strength is not required. Rolled sections (angles, channels, etc.) are not available in HPS grades.

4. The fewer the girders, the greater the economy. Girder spacing must be compati­ble with deck design, but sometimes other factors, such as maintaining traffic dur­ing a future deck replacement, govern selection of girder spacing. For economy, girder spacing should be 10 ft (3 m) or more.

5. Transverse web stiffeners, except those serving as diaphragm or cross-frame con­nections, should be placed on only one side of a web.

6. Web depth may be several inches larger or smaller than the optimum without sig­nificant cost penalty.

7. A plate girder with a nominally stiffened web—1/16 in (1.6 mm) thinner than an unstiffened web—will be the least costly or very close to it. (Unstiffened webs are generally the most cost-effective for web depths less than 52 in (1320 mm). Nominally stiffened webs are most economical in the 52- to 72-in (1320- to 1830-mm) range. For greater depths, fully stiffened webs may be the most cost-effective.)

8. Web thickness should be changed only where splices occur. (Use standard – plate-thickness increments of 1/16 in (1.6 mm) for plates up to 2 in (51 mm) thick and 1/8-in (3.2-mm) increments for plates over 2 in (51 mm) thick.)

9. Longitudinal stiffeners should be considered for plate girders only for spans over 300 ft (92 m).

10. Not more than three plates should be butt-spliced to form the flanges of field sec­tions up to 130 ft (40 m) long. In some cases, it is advisable to extend a single flange-plate size the full length of a field section.

11. To justify a welded flange splice, about 700 lb (318 kg) of flange steel would have to be eliminated. However, quenched-and-tempered plates are limited to 50-ft (15-m) lengths.

12. A constant flange width should be used between flange field splices. [Flange widths should be selected in 1-in (25-mm) increments.]

13. For most conventional cross sections, haunched girders are not advantageous for spans under 400 ft (122 m).

14. Bottom lateral bracing should be omitted where permitted by AASHTO specifica­tions. Omit intermediate cross-frames where permitted by AASHTO, but indicate on the plans where temporary bracing will be required for girder stability during erection and deck placement. Space permanent intermediate cross frames, if required, at the maximum spacing consistent with final loading conditions.

15. Elastomeric bearings are preferable to custom-fabricated steel bearings.

16. Composite construction may be advantageous in negative moment regions of composite girders.

Designers should bear in mind that such techniques as finite-element analysis, use of

high-strength steels, and load-and-resistance-factor design often lead to better designs.

Consideration should be given to use of 40-in-deep (1016-mm) and 44-in-deep (1118-mm) rolled sections. These may be cost-effective alternatives to welded girders for spans up to 100 ft (30 m) or longer. Economy with these beams may be improved with end-bolted cover-plate details. Equally important is the availability of material, either in the form of rolled beams or plates. Long-lead items may cause schedule delays and contractor claims, which increase the cost of construction. Contract documents that allow either rolled beams or welded girders ensure cost- effective alternatives for owners.

With fabricated girders, designers should ensure that flanges are wide enough to provide lateral stability for the girders during fabrication and erection. Flange width should be at least 12 in (305 mm), but possibly even greater for deeper girders. The AISC recommends that, for shipping, handling, and erection, the ratio of length to width of compression flanges should be about 85 or less.

Designers also should avoid specifying thin flanges that make fabrication difficult. A thin flange is subject to excessive warping during welding of a web to the flange. To reduce warping, a flange should be at least 3/4 in (19 mm) thick.

To minimize fabrication and deck-forming costs when changes in the area of the top flange are required, the width should be held constant and required changes made by thickness transitions.

To get cost-effective results from the many different designs of fabricated girders that can satisfy the requirements of specifications, designers should obtain advice from fabricators and contractors whenever possible.

The welded steel plate girder bridge (Fig. 4.9) extends the span range of deck-type bridges (bridges having all the structural support below the deck slab) well beyond the range of rolled steel beams or precast prestressed-concrete beams.

Whereas haunched girders were economical in the past for long spans, the current practice, strongly advocated by the steel-fabricating industry, is to use parallel-flange girders wherever possible. This is an economic consideration rather than an aesthetic one. Properly configured haunched girders are thought by many to be more pleasing. They permit a shallower structure depth at mid span, which can result in a lower grade line and consequent savings in roadway construction cost. However, parallel-flange girders can be fabricated more rapidly and economically than haunched girders. As an example of long-span parallel-flange steel girder construction, the Tennessee DOT has designed a continuous 1717-ft-long (523-m) parallel-flange five-span steel plate girder bridge of girder-floor beam-stringer type construction, including two spans of 460 ft (140 m) each. This design uses ASTM A36 and A572 steel in the webs, and A572 and A517 steel in the flanges. The A517 steel, which has a minimum yield strength of 100 kip/in2 (690 MPa), is used for the flange plates at points of maximum stress over the piers and at mid span of one of the 460-ft (523-m) spans.

In designing a steel plate girder bridge for economy, designing for minimum weight does not always result in the most economical girder. The cost saved by reducing web or flange plate width or thickness may be more than offset by the cost of making the welded splices. Cost data should be obtained from local fabricators to make this comparison. One rule of thumb is that the weight saved by a change of flange plate thickness should be at least 1500 lb (680 kg). Also, it is generally desirable to use a constant-width flange plate to reduce fabrication and construction costs.

The use of excessively thin webs and narrow flanges, while saving weight, can result in flimsy sections that require special handling and erection equipment such as strong­backs. If such measures are not employed, the girder may be damaged in handling. Either consequence may more than offset the cost saved through weight reduction. For this rea­son many states have adopted minimum plate dimensions that are greater than minimum requirements of AASHTO or industry recommendations.

The steel-beam bridge uses rolled steel beams as shown in Fig. 4.8. Beam depths of 44, 40, and 36 in (1118, 1016, and 914 mm) are available, as well as shallower sections. Check with producers on current availability of the deeper sections from domestic sources because federal law applicable to federally aided projects, as well as many state laws, prohibits the use of foreign steel.

Steel beams may be made continuous by welding or bolting sections in the field. In the past, some states made welded connections at the piers, and currently at least

one state makes welded connections at contraflexure points, supporting the field sections temporarily and providing enclosures to shield the joint from wind. More commonly, field sections are spliced by high-strength bolts, using web-and-flange splice plates. Bolts may be installed using calibrated wrenches, by the turn-of-nut method, or by use of tension-indicating washers, depending on what the designer allows and what the erector prefers to use. With all methods of bolting, it is important to use a procedure and sequence of bolting that will compact the joint and prevent a bolt initially adequately tightened from losing tension when subse­quent bolts in the joint are tightened. Fasteners are generally ASTM A325 or A490 high-strength bolts.

To increase the span capacity of a rolled beam, or to permit a lighter beam to be used, cover plates may be added above the top flange and below the bottom flange in regions of high bending stress due to both positive and negative moments. The fatigue strength at the end of the cover plates, which is generally at a point of low maximum stress but high stress range, is much less than the fatigue strength of the unplated beam. Allowable fatigue stresses must not be exceeded, and this consideration may favor an unplated beam. However, an improved detail is available that uses bolts at the end of the plate, and the fatigue strength is somewhat higher. New Jersey DOT requires full-length cover plates, with termination about 2 ft (610 mm) from the end of the beam where the stress range is very small.

Prestressed-concrete beams of the basic I-shape, but with variations, can be used over approximately the same range of spans as steel beams. The deepest AASHTO standard prestressed beams (72 in or 1828 mm) have a somewhat greater simple-span capacity than 36-in-deep (914-mm) rolled steel beams, although deeper rolled beams are avail­able. This type of bridge is illustrated in Fig. 4.7.

Prestressed-concrete beams are heavier to transport and erect than steel beams, and require more care in handling. A prestressed-concrete beam can be destroyed if it is not maintained in an upright position.

I-beams may be standard AASHTO-PCI sections or conform to individual state standards. Depth varies from 28 in (711 mm) for the little-used AASHTO type I to 72 in (1828 mm) for the AASHTO type VI and BT-72 (1828 mm) bulb-tee. The basic differ­ence between the AASHTO type V and type VI beams and the bulb-tee beams, all of which have 3.5-ft-wide (1067-mm) top flanges, is that the bulb-tees have a thinner web (6 in instead of 8 in or 152 mm instead of 203 mm) and shallower top and bottom flanges. The bulb-tees have a flatter slope on the top of the bottom flange, as well. A variant of both is the modified AASHTO type VI, which uses the side forms for the AASHTO type VI beam but only a 6-in (152-mm) web. Individual analysis will determine which shape is best, but only shapes that are available from local precasters should be investigated unless the project is large enough to economically justify the purchase of special forms. Sometimes, bulb-tee sections are modified to have deeper web sections to increase their capacity, hence the span length.

As with prestressed-concrete box-beam bridges, the prestressing strands may include deflected or debonded strands. When strands are deflected and a number of beams are cast in line on a casting bed, resulting in many hold-down or hold-up points, stressing procedures should be used and verified that limit the maximum prestress loss due to friction to the amount permitted by specifications.

For very long bridges with repetitive spans over water, and where there is a precasting plant at a site from which the bridge units can be delivered by barge, the option of pre­cast deck units consisting of the beams, diaphragms, and deck slab cast monolithically should be considered.