Category HIGHWAY ENGINEERING HANDBOOK

There are also a number of prefabricated modular wall systems in use. Such systems are generally composed of modules or bins filled with soil, and function much like gravity retaining walls. The bins may be of concrete or steel, and can be used in most cases where conventional gravity, cantilever, or other wall systems are considered. AASHTO indicates that such walls should not be used on curves less than 800 ft in radius, unless a series of chords can be substituted; or where the calculated longitudinal differential set­tlement along the face of the wall is excessive. Also, durability considerations must be addressed, particularly where acidic water or deicing spray is anticipated.

8.9 MSE BRIDGE ABUTMENT WALLS

The abutment wall is an earth-retaining wall supporting traffic surcharge load and heavy loads from the bridge superstructure. The geosynthetic-reinforced soil (GRS) earth-retaining wall is a subset of the MSE wall. The technology of GRS systems has been used extensively in transportation systems including earth-retaining walls, road­way pavement subgrades, and foundation improvement for heavy traffic loads such as bridge abutments. The increasing use and acceptance of geosynthetic soil reinforce­ment has been triggered by a number of factors, including cost savings, aesthetics, simple and fast construction techniques, and excellent performance. A comparatively new application of this technology is the use of GRS in bridge abutments and roadway approaches. When compared to conventional bridge substructures involving the use of deep foundations to support bridge abutments, the use of geosynthetic-reinforced soil systems has the potential for alleviating the “bump at the bridge” problem caused by differential settlements between the bridge abutment and the approaching roadway. It is especially effective where the GRS can be extended beyond the typical rectangular reinforcing zone of the wall and truncated gradually into a trapezoidal reinforcing zone toward the approach roadway. The FHWA published preliminary design details for bridge superstructures directly supported by MSE walls with panel facings and steel reinforcements in 1997 (Elias and Christopher, 1997), and it was included in the AASHTO 1998 Standard Specifications for Highway Bridges. A recently published FHWA report (FHWA, 2000) describes three studies on GRS bridge-supporting structures: a load test of the Turner-Fairbank pier in McLean, Virginia, in 1996; a load test of the Havana Yard piers and abutment in Denver, Colorado, in 1996-1997; and a study of a production bridge abutment constructed in Black Hawk, Colorado, in 1997. These studies have demonstrated excellent performances with negligible creep deformations of GRS bridge-supporting structures constructed with closely spaced reinforcements and well-compacted granular backfill. The maximum surcharge pressure was 4.2 kip/ft2 (200 kPa). This FHWA report concluded that the GRS abutments are clearly viable and adequate alternatives to bridge abutments supported by deep foundations or by metallic reinforced soil abutments. A complete literature overview of studies on GRS structures supporting high-surcharge loads has been presented (Abu-Hejleh et al., 2000).

The most prominent GRS abutment for bridge support in the United States is the new Founders/Meadows Parkway structure, located 20 mi south of Denver, Colorado, which carries Colorado State Highway 86 over U. S. Interstate 25 (Fig. 8.60). This is the first major bridge in the United States built on spread footings supported by a concrete block facing geosynthetic-reinforced soil system, eliminating the use of traditional deep foundations (piles and caissons) altogether. Figure 8.61 shows the bridge super­structure supported by the “front GRS wall,” which extends around a 90° curved corner in a “lower GRS wall” that supports a “concrete wing wall” and a second-tier “upper GRS wall.” Figure 8.62 shows a plan view of the completed two-span bridge and approaching roadway structures. Each span of the new bridge is 113 ft (34.5 m) long and 113 ft (34.5 m) wide, with 20 side-by-side, precast, prestressed, concrete box girders. There are three monitored cross-sections (sections 200, 400, and 800) along the faces of the “front GRS wall” and “abutment wall.” Figure 8.63 depicts a typical monitored cross-section with various wall components and drainage features in the backfill. To keep the water out of the GRS, several drainage systems were used in the trape­zoidal extended reinforcing zone, including an impervious membrane and collecting drain at the top and a drainage blanket and pipe drain near the toe of the embankment cut slope. Figure 8.63 also illustrates how the bridge superstructure loads (from bridge deck to girders) are transmitted through the girder seat to a shallow strip footing placed directly on the top of a geogrid-reinforced concrete block facing earth retaining wall. The centerline of the abutment bearing and edge of the footer are located 10 ft (3.1 m) and 4.4 ft (1.35 m), respectively, from the facing of the front GRS wall. This short reinforced-concrete abutment wall supports the bridge superstructure, including two winged walls cantilevered from the abutment. This wall, with a continuous neoprene sliding and bearing interface at the bottom, rests on the center of the spread footing.

Bridge Structure

East

FIGURE 8.62 Plan view of the Founders/Meadows structure showing the locations of monitored sections (200, 400, and 800) and construction phases (I and II). (From Research Report, Colorado Department of Transportation, Denver, Colo., with permission)

It confines the reinforced backfill soil behind the bridge abutment (see Figs. 8.61 and 8.63) and also partially supports the bridge approach slab. The old bridge was maintained in service during construction of the new bridge. A phased construction of the almost 9-m-high (29-ft), two-tier, U-shaped abutments began in July 1998, and the bridge was completed in just 12 months. This included a temporary bracing at the south side of the replaced bridge and a temporary GRS wall at the north of phase 1 of the partial new bridge. The Colorado Department of Transportation (CDOT) designed the Founders/Meadows structure in 1996, 1 year before the FHWA report of Elias and Christopher (1997) was published. It was expected that water could be kept out of the claystone bedrock formation below the base of the reinforced backfill, and that the use of an extended reinforced transition zone (Fig. 8.63) would lead to adequate overall stability for the structure and minimize settlements of the GRS wall system.

Several of the common causes for development of bridge bumps were avoided or eliminated completely in the design of the Founders/Meadows structure. The main cause of uneven settlements in typical systems is the use of different foundation types. While the approaching roadway structure is constructed on compacted backfill, deep foundations such as steel H piles typically are adopted to transfer the heavy bridge abutment loads to bedrock or competent bearing strata. At the Founders/Meadows structure, in order to minimize uneven settlements between the bridge and the approaching roadway, the approach embankment and the bridge abutment wall backfills were further integrated into an extended reinforced soil zone (Fig. 8.63). A second cause of differential settlements is erosion of the fill material behind and around the abutment wall induced by the surface water runoff collected from the bridge deck.

Several measures were implemented to prevent the peculation of surface water, as well as intercepted groundwater, from reaching the reinforced soil mass and the bedrock at the base of the fill (e. g., placement of impervious membranes with collector pipes as shown in Fig. 8.63). Finally, a third potential cause of differential settlements is sea­sonal temperature changes, which may induce expansion and contraction of the bridge superstructure with the abutment back wall pushed against the reinforced fills. A 3-in-thick (75-mm) low-density, compressible expanded polystyrene sheet spacer was installed between the reinforced backfill and the abutment back walls (see Fig. 8.63). It was expected that this system would isolate the thermally induced cyclic movements of the bridge superstructure without affecting the retained reinforced backfill.

The performance of bridge structures supported by GRS abutments had not been tested under actual service conditions to merit acceptance without reservation in highway construction. Consequently, the Founders/Meadows structure was considered experi­mental and comprehensive material testing, instrumentation, and monitoring programs were incorporated into the construction operations. Three sections of the GRS system were instrumented to provide information on the structure movements, soil stresses, geogrid strains, and moisture content during construction and even after opening the structure to traffic. The results gathered from the instrumentation program verified the suitability of CDOT and AASHTO design procedures and assumptions regarding the use of GRS walls to support bridge abutments and as a measure to alleviate the bridge bump problem. Several CDOT research reports present a summary, analysis, and assessment of all the movement results of the Founders/Meadows structure, collected at various construction stages and while the structure was in service for approximately 35 months after opening to traffic. These results include displacements of the front wall facing, settlement of the bridge footing, and differential settlements between the bridge and approaching roadway structures. These reports also provide recommendations for future design and construction of GRS abutments directly supporting bridge and approaching roadway structures. The unique features of the Founders/Meadows Bridge, and other perceived advantages of GRS walls commonly used in Colorado, as well as the excellent performance of full-scale GRS abutments and piers in monitoring programs, convinced CDOT engineers of the feasibility of GRS earth-retaining walls and abutments.

Evaluation of a global safety factor that includes the nailed soil and the surrounding ground requires determination of the critical sliding surface. This surface may be located totally inside, totally outside, or part inside and part outside the nailed zone. Limit equilibrium methods are usually used, and the Davis method is recommended because of its simplicity and availability in the public domain (C. K. Shen et al., “Field Measurements of an Earth Support System,” Journal of the Geotechnical Division, American Society of Civil Engineers, vol. 107, no. 12, 1981). The Davis method has been modified (V. Elias and I. Juran, “Soil Nailing,” Report for FHWA, DTFH 61-85-C, 1988) to permit input of interface limit lateral shear forces obtained from pullout tests, separate geometric and strength data for each nail, facing inclination, and a ground slope at the top of the wall. The concrete facing elements (shotcrete, cast-in-place concrete, or prefabricated panels) are considered for design to be analogous to a beam or raft of a unit width equal to the nail spacing supported by the nails.

8.8.4 Contracting Practices

Although procurement and contracting practices vary among the European countries, there are some common elements that tend to distinguish European practices from those in the United States. These include (1) strong industry, academic, and government
cooperation in research and development and the introduction of new technologies; (2) a partnering approach among all parties involved in a particular project; (3) less litigation; and (4) a high level of contractor involvement in the conceptualization and design phases, as well as during construction.

In France, the contractor design-build approach appears to be dominant. For public agency work, a prequalified group of contractors are typically asked to prepare a final design and bid, based on a preliminary design prepared by the owner or the owner’s consultant. Alternative designs may also be prepared by the contractor at this time, and may be selected if they are technically and financially viable and meet the overall performance and scheduling requirements of the project. French contractors tend to be much larger and stronger than their U. S. counterparts, and the major groups tend to support significant research and development efforts. Contractor-consultant-academic – government cooperation in areas requiring major research and development is particularly well developed in France.

In Germany, public agency work is again usually bid on a conceptual or prelimi­nary design prepared by or for the agency, with the contractor required to submit a bid on the original design and also encouraged to submit any alternative design that will provide an equivalent wall at a reduced price. Ultimately, award is made for the lowest – cost responsive bid. Soil nailing in Germany requires the involvement of one of a small group of prequalified or “licensed” contractor organizations. As in France, these contractors tend to be technically and financially very strong. Private work, like public work, tends to be awarded on the basis of low bid.

Based on the European, and particularly the French, experience, two main recommen­dations are offered for encouraging the development of innovative construction methods and improving the construction performance for such methods. First, stronger and more formal government-academic-industry cooperation should be established to develop new technologies and disseminate the information in a nonproprietary manner. This should also include participating in corresponding European programs when the opportunity arises and when the information will be of mutual benefit. Second, alternative bidding, including contractor design-build alternatives, performance-oriented specifications, and the use of carefully prequalified specialty contractors, should be encouraged.

The following typical details and design procedure are based primarily on Caltrans’ method for use on highway construction, but the method is very similar to other methods presently in practice. Design facing pressures are based on the French Clouterre empirical method. The cast-in-place portion of the facing is designed for this pressure for permanent

FIGURE 8.57 Section through facing of soil nailed wall showing concrete reinforcement and soil nail connection. (From J. W. Keeley, Soil Nail Wall Facing: Sample Design Calculations, Federal Highway Administration, 1993, with permission)

walls only. The strength of the shotcrete construction facing is ignored. Only the ultimate limit state is addressed; no serviceability calculations are made for cracking or deflections. Sample design calculations are illustrated following the presentation of the procedure.

Typical Details. (See Figs. 8.57 to 8.59.)

1. Use a shotcrete layer with a 4-in (100-mm) minimum thickness.

2. Include a single layer of welded wire fabric at mid-thickness. Common options are:

6 X 6-W4.0 X W4.0 (4 gauge wire; diameter, 0.225 in; cross-section area = 0.080 in2/ft or 0.17 mm2/mm)

4 X 4-W2.9 X W2.9 (6 gauge wire; diameter, 0.192 in; cross-section area = 0.087 in2/ft or 0.18 mm2/mm)

3. Use two continuous horizontal no. 4 grade 60 reinforcing steel bars at each nail.

4. Use 1-ft-wide (300-mm) vertical geocomposite drain between nails; connect the geo­composite drain to a 2-in-round (600-mm) plastic weep hole outlet drain just above finished grade near the bottom of the wall.

5. Place the ASTM A36 steel nail bearing plate, 1 in X 9 in X 9 in (25 X 225 X 225 mm) with wedge washer and nut on the outside face of the shotcrete; set into place before the shotcrete hardens. Add studs to this plate to engage permanent facings that are placed over this initial shotcrete layer.

Step 1: 4-in (100-mm) Shotcrete Construction Facing. This is the only facing required for temporary walls (service life less than 18 months) and the first portion required for permanent walls. It is placed immediately after each stage of excavation and nail placement. Current AASHTO and American Concrete Institute (ACI) codes do not address the loadings or the structural capacities for this facing. Therefore, many current designs rely on details that have shown good performance on previous projects rather than design calculations.

Step 2: 8-in (200-mm) CIP Permanent Facing—Compute Design Nail Load and Pressure at Facing. The design nail load at the facing is computed for the given nail size, steel grade, and nail spacing according to the French Clouterre empirical method. The French determined through field tests that the nail load at the facing (T0) did not exceed about one-half the maximum nail load (T ) near the soil failure surface. They

v max

established a design nail load for the facing that varies from 0.6 times T for closely spaced nails to 1.0 times T for larger nail spacings. T is the ultimate limit state established for the nail tension (Caltrans procedure) at 0.75 X f (T = A. X 0.75 X

у v max nail

fy), where fy is the yield strength of the nail. The design pressure for the facing is then simply the design nail load at the facing (T0) divided by the nail tributary facing area (i. e., horizontal nail spacing times vertical nail spacing).

Step 3: 8-in CIP Permanent Facing—Design for Flexure. The cast-in-place facing is designed so that its ultimate strength is greater than the moments in the facing computed by simple continuous beam equations for the facing pressure from T0. Only one layer of grade 60 reinforcing steel placed near the middle of the section is used. (See Fig. 8.57.) The controlling d is used for the ultimate strength computation.

Step 4: 8-in CIP Permanent Facing—Nail Connection Design. The connection of the nail to the cast-in-place facing is designed to carry the nail’s ultimate limit state in tension, T. The nail bearing plate is sized to limit the bearing pressure from T to the

max max

ultimate value allowed by AASHTO (0.6f’c). The plate thickness is determined to provide sufficient bending strength for the moment from the bearing pressure about the nail nut. Studs are welded to the bearing plate to carry Tmax entirely by the 8-in (200-mm) CIP per­manent facing. The ultimate punching shear capacity of the steel embedment is computed according to American Concrete Institute specifications.

Sample Design Calculations for Soil Nail Wall Facing (Based on Caltrans Methods)

Step 1: 4-in Shotcrete Construction Facing. Details as previously described may be used.

Step 2: 8-in CIP Permanent Facing—Concrete Design Nail Load and Pressure at Facing. Given No. 8 nails; f = 60 kips/in2; Anail = 0.79 in2; horizontal nail spacing Sh = 6 ft; and vertical nail spacing Sv = 6 ft.

Begin by calculating the design nail loads.

T = design nail load at soil failure surface at ultimate limit state

max

= 0.75f A = 0.75 X 60 X 0.79 = 35.6 kips

у nail

T0 = design nail load at facing at ultimate limit state

(Note: This is the French Clouterre equation.)

The design nail load T0 is then used to calculate the facing design pressure.

Wu = facing design pressure at ultimate limit state

27.0

——- = 0.75 kip/ft2

6 X 6

Step 3: 8-in CIP Permanent Facing—Design for Flexure. Note: Try this wall section with No. 6 reinforcing bars spaced at 12 in horizontally and vertically; f c = 3250 lb/in2. c

Required ultimate moment per foot (horizontal and vertical; positive and negative)

Ultimate design capacity for trial section

8-in CIP permanent facing

depth of concrete compression

фмп=Фм(^ – _

= 0.9 X 0.44 X 60 X ^3.38 – __ j X _ = 5.9 ft • kips/ft > 2.7 OK

Check of minimum steel requirements. According to AASHTO, the tension rein­forcement must be equal to or greater than the lesser of that required to develop a moment (1) 1.2 times the cracking moment (based on the gross section modulus S ) and (2) 1.33 times that required by analysis for the specified loading conditions. This leads to the following equations for the cross-section area of the reinforcement:

1.2 X S X 7.5V2

_______ g_______ _

ф__ X 0.9d

The strength reduction factor ф is 0.90. Substitution gives the following results:

Thus, As must be at least 0.26 in2/ft to meet these requirements. For temperature and shrinkage, minimum As is 0.13 in2/ft (No. 4 @18 in). The final selection is No. 6 rein­forcing bars spaced at 12 in, which provides As of 0.44 in2/ft, a nominal level of reinforc­ing. For complete designs, also check cantilever sections at the top and bottom of the wall and any other special facing sections, such as at expansion or contraction joints.

Step 4: 8-in CIP Permanent Facing—Nail Connection Design

Design bearing plate at ultimate limit state for Tmax

T = 0.75f A = 0.75 X 60 X 0.79 = 35.6 kips

max y nail

Calculate the ultimate concrete bearing strength under the plate using a strength reduction factor of ф = 0.70. Therefore,

Ultimate concrete bearing strength = 0.85фf’c

= 0.60f’ .

c min

where w = ultimate bearing pressure.

Required steel area for studs to resist Tmax. The design strength of the stud fds is determined as the lesser of (1) the stud yield strength (f = 50 kips/in2) multiplied by a strength reduction factor of ф = 0.90 and (2) 80 percent of the stud tensile strength (fu = 60 kips/in2). Therefore,

fds = min Щ; 0.8fJ ф^ = 0.9 X 50 = 45 kips/in2

0. 8fu = 0.8 X 60 = 48 kips/in2

Therefore

fds = 45 kips/in2 T

Required stud area As = —

fds 35.6

Try 4/2-in ф studs.

As = 4 X 0.196 = 0.79 in2 OK

Check of anchor head bearing for ‘A-in ф stud. ds = 0.5 in. Determine head area Ah.

dh^2

Ah 0.79

A 0.196

T. = 0.312 in > 0.25 in

Head thickness is OK.

Ultimate connection embedment design capacity Pd Pd = 4ф VfA

d c cp

f’ = 3250 lb/in2

c

ф = 0.65 = strength reduction factor in this case
Acp = 291.7 in2 = effective stress area (see Fig. 8.59)

Pd = 4 X 0.65 X V3250 X 291.7 X = 43.2 kips

P, > T = 35.6 kips OK

d max

The construction sequence is typically to excavate, nail, and shotcrete the face in increments from the top down. Figure 8.56 shows a schematic of a sequence for underpass widening. Where face stability is a concern, a flashcoat of shotcrete may be applied before nail installation. The most common method of nail installation in

FIGURE 8.56 Construction sequence for soil nailed wall used in widening of underpass. (From Oregon Department of Transportation, with permission)

STEP 7: Place second structural application to full height and architectural finish

FIGURE 8.56 (Continued)

Europe, as in the United States, is the drill-and-grout method. Most commonly, the steel tendon is installed prior to grouting, although this sequence is sometimes reversed. In France, however, very significant use is also made of driven nails without grouting. Other specialty techniques for installing nails include jet grout nailing (France), driven nails with an oversize head and subsequent grouting of the annulus (Germany), and compressed-air explosive injection of nails (United Kingdom).

Small hydraulic, track-mounted drill rigs of the rotary-percussive type are most com­monly used to install nails. These rigs can work in relatively confined surroundings and are therefore compatible with many of the constraints associated with crowded urban

environments. Open-hole drilling methods are predominant, with cased-hole methods used in particularly difficult ground conditions. The most common grouting method used with open-hole drilling is the low-pressure tremie method. Where extensive use of casing is required, alternative methods of construction are often more cost-competitive.

Steel tendons typically used for drill-and-grout soil nails usually consist of /4- to 2-in­diameter (20- to 50-mm) bars with a yield strength in the range of 60 to 70 kips/in2 (420 to 500 N/mm2). These steels exhibit ductile behavior under bending action. The driven nails used commonly in France are typically steel angle sections, which show better ability to deal with subsurface obstructions such as cobbles and small boulders than do circular steel sections.

Drainage is a critical aspect of soil nail wall construction. Face drainage is virtually always used with permanent walls, and very commonly used with temporary walls. Face drainage usually consists of synthetic drainage elements placed between the shotcrete and the retained soil, and may be typically 8- to 12-in-wide (200- to 300-mm) synthetic strips or perforated pipes. Depending on the site groundwater conditions, face drainage may be supplemented with weep holes through the facing and longer horizontal perforated drain pipes. Control of surface water is also an important element of the drainage system.

Temporary soil nail wall facings generally consist of shotcrete 3 to 4 in (80 to 100 mm) thick and a single layer of wire mesh. Permanent shotcrete walls 6 to 10 in (150 to 250 mm) thick are very common in Germany, and these walls typically include a second layer of wire mesh. For architectural reasons, permanent walls of precast panels and cast-in­place concrete are also commonly used in France and Germany.

Testing and monitoring during construction are an important aspect of soil nail wall construction in Europe. Nail bond testing is almost universally performed, to confirm the assumptions made during design or to enable redesign in the event that the design assumptions cannot be realized. For relatively homogeneous sites, typically 3 to 5 percent of the nails will be tested, depending on the size of the job. Testing is also undertaken whenever changed geologic or construction conditions occur. Wall performance moni­toring usually consists of measuring horizontal wall movements during construction. Some contractors make more use of inclinometers for displacement monitoring. Maximum horizontal displacements are typically in the range of 0.1 to 0.3 percent of the height of the wall, depending on ground conditions. Strain gauging of nails, together with the use of load cells at the nail head, is usually reserved for experimental walls.

The level of quality assurance and control monitoring varies significantly. In Germany, for example, the QA-QC inspector might be on the job from 10 percent to full time.

Overall, soil nail wall performance in Europe has been very good. Problems during construction have typically been associated with encountering loose fill, granular soil with no apparent cohesion, water, and constructed obstructions such as utility trenches. Other problems have been associated with a contractor’s failing to construct the wall in accordance with the plans and specifications (e. g., eliminating nails, overexcavation of lifts). Frost action on fully bonded nails has also resulted in development of very large loads near the head of the nail, where no insulating protection has been provided in the wall design.

The stability of a soil nailed structure relies on (1) transfer of resisting tensile forces generated in the inclusions in the active zone into the ground in the resistant zone, through friction or adhesion mobilized at the soil-nail interface, and (2) passive resistance developed against the face of the nail. Ground nailing using closely spaced inclusions produces a composite coherent material. As shown in Fig. 8.54, the tensile forces gen­erated in the nails are considerably greater than those transmitted to the facing.

The design procedure for a nailed retaining structure includes (1) estimation of nail forces and location of the potential sliding surface, (2) selection of the reinforcement type, cross-sectional area, length, inclination, and spacing, and (3) verification that stability is maintained during and after excavation with an adequate factor of safety. Methods for determining tensile, bending, and shear stresses in the nails are given by FHWA based on a limit equilibrium analysis.

The majority of soil nailed retaining structures constructed in France are based on two distinct technologies: (1) the method of Hurpin, with nails driven into the ground on close spacing, i. e., vertical and horizontal spacing equal to or less than 3 ft (1 m), and (2) widely spaced grouted nails. With the method of Hurpin, the nails (generally reinforcing bars) are relatively short and are driven into the ground by percussion or vibratory methods. The relatively high nail density permits thinner wall facings. In walls with widely spaced nails, the nails are generally longer. Typical data for soil nailed walls with a vertical facing and a horizontal earth pressure are shown in Table 8.11. The “nailing density” listed in Table 8.11 is a dimensionless parameter representing soil nails placed in a uniform pattern. It is defined as

(8.25) where T = ultimate tensile force that can be mobilized at head of nail

Г

Sh = horizontal spacing between nails Sv = vertical spacing between nails L = length of nails у = total unit weight of soil

This parameter represents the maximum tensile force in a nail as it relates to the weight of the soil reinforced with a chosen grid spacing.

A full set of preliminary design charts is included in the FHWA translation of Recommendations Clouterre, 1991. Diagrams for an angle of installation of the nails of i = 20° are shown in Fig. 8.55 for illustration.

Figure 8.55 provides a preliminary chart for a soil nailed wall. It seeks to define in approximate terms the lengths, spacings, and resistance values of the nails to ensure

internal and external stability. It may be used in an early evaluation stage based on macro assumptions such as homogeneous soil, identical and evenly spaced loads in the nails, and pure tension in the nails; the bending stiffness of the nails is neglected, regardless of the angle of incidence of the potential failure surface. This approach is based on the classic method of vertical slices with circular potential failure surfaces. The charts are based on a system of coordinates that characterize the shear resistance of the soil.

c 20

N = ————— = —————————- = 0.10 tan ф = tan 35° = 0.70

7H 20 X 10 v

Draw a line from the origin O to M. The safety factor F is the ratio OM/OA. Therefore, for a safety factor of 3/2, locate point A two-thirds of the distance along the line OM. Interpolation gives the required nailing density d as 0.33. Thus:

TL

7—

0. 33 X 20 X 0.8 X 10 = 52.8 kPa

Thus, for a nail tensile force TL, the spacings Sh and Sv can be determined. The result from the chart should be generally conservative and used only for preliminary evaluation.

The final design for stability of a soil nailed wall is analyzed either by calculating the deformations or by using limit equilibrium design. The first method uses finite ele­ment calculation and has not been refined to the point where there is an “acceptable” procedure. In Europe to date, there has been considerable diversity in some details among the various design approaches, both within and across national boundaries. A most significant factor is the postulated mechanism by which nails are considered to reinforce a soil mass. For nails installed nearly parallel to the direction of maximum soil tensile strain (e. g., near-horizontal nails and a near-vertical excavation face), the prevailing opinion is that the reinforcing action is predominantly related to tensile loading within the nails. Under service load conditions, the contribution of shear or bending is considered negligible. As failure conditions are approached, the contribution of shear or bending action is more significant but still small. From a practical point of view, however, it is recognized that the soil nails should exhibit ductile behavior in response to bending in order to minimize the potential for sudden failures related to brittleness. Where reinforcing elements are used as dowels and are oriented nearly perpendicular to the direction of maximum shear strain, the shearing, bending, and tensile action of the reinforcement should be considered.

All design methods are based on concepts of limiting equilibrium or ultimate limit states. Various types of potential slip surface are considered, including circular, log spiral, and bilinear wedge. In general, each of the methods appears to provide a satis­factory representation for design purposes. Consistent with the above, most design computer codes consider only the tensile action of the nails, but some also permit con­sideration of the shear or bending action of the nails. Almost all of the design methods do not explicitly consider the potential for pullout of the reinforcing nails within the active block between the facing and the slip surface. It is implicitly assumed that the nail – soil adhesion within this zone, together with the structural capacity of the facing, will be sufficient to prevent this type of failure. Some design approaches offer strict guide­lines for the required structural wall capacity to prevent such active-zone failures, but others appear to rely on experience and do not directly address this issue.

On the basis of the overall reinforcing requirements determined from the limiting equilibrium design calculations, the reinforcing steel is empirically proportioned. In general, designers use nails of uniform length and cross-sectional area, on a uni­form spacing. For drilled and grouted nails, the nail spacing is typically in the 3- to 6-ft (1- to 2-m) range. For driven nails, much higher densities (typically 1.5 to 2 nails per
square meter) are used. The nail lengths are typically in the range of 60 to 80 percent of the height of the wall, but may be shorter in very competent rocklike materials and longer for heavy surface surcharge or high seismic or other operational loading.

As noted above, facing design requirements are empirically determined using a variety of techniques. German practice requires the use of a uniform facing pressure equivalent to 75 to 85 percent of the active Coulomb loading. The Clouterre recom­mendations require designing the facing and connectors to support between 60 and 100 percent of the maximum nail loading (for both ultimate and serviceability limit states), depending on the nail spacing.

Almost all shotcrete failures in slope stabilization applications have resulted from

inadequate drainage. Therefore, drainage is a critical design and construction element.

Drainage from behind the shotcrete face can be provided by the following methods:

1. Surface interceptor ditch. Excavate a shallow ditch along the crest of the excavation to lead away surface water. Drainage gutters or lined ditches are recommended immediately behind the top of the wall.

2. Prefab geotextile drains. Place 12-in-wide (300-mm) prefabricated geotextile drain strips (Miradrain 6000, Amerdrain 200, etc.) vertically prior to applying the shotcrete. Typical spacings are the same as the horizontal nail spacing. Extend the drain mats down the full height of the excavation and discharge into a collector pipe at the base.

3. Weep holes and horizontal drains. Install 2-in-diameter (50-mm) PVC pipe weep holes on approximately 10-ft (3-m) centers through the shotcrete face where heavier seepage is encountered. Plug the pipe temporarily when shotcrete is applied. Longer PVC horizontal drain pipes can also be installed in heavy seepage areas.

8.8.3 Wall Facing Systems

Temporary walls are typically left with a rough shotcrete face—“gun” finish—with weep holes and protruding nail heads. For permanent walls, where the rough finish is aesthetically unacceptable, the following face options are available:

1. Separate fascia wall. As an alternative to the exposed shotcrete finish, the shotcrete can be covered with a separate concrete fascia wall, either cast in place (CIP) or constructed of precast panels. The CIP section is typically a minimum of 6 to 8 in (150 to 200 mm) thick. Precast face panels can be smaller modular panels or full – height fascia panels such as those used to cover permanent slurry walls. A disadvantage of the smaller modular face panels is difficulty of attaching the face panels to the nail heads and some proprietary patent restrictions. A disadvantage of full-height precast panels is that, because of practical constructibility weight and handling limitations, their use is limited to wall heights less than 25 ft (8 m).

2. Permanent exposed shotcrete facing. Present technology for shotcrete place­ment is such that the final shotcrete layer can be controlled to close tolerances, and with nominal hand finishing, an appearance similar to a CIP wall can be obtained (if desired). The shotcrete, whether left in the natural gun finish or hand-textured, can also be colored either by adding coloring agent to the mix or by applying a pigmented sealer or stain over the shotcrete surface. Only experienced, well-qualified structural shotcrete specialty contractors should place and finish the permanent structural shot – crete. “Wet-mix” shotcrete should be specified instead of dry-mix because good quality control is easier with wet-mix. Also, wet-mix shotcrete can be air-entrained for improved freeze-thaw durability, whereas dry-mix cannot.

Soil nailed walls have some similarities with MSE walls but also some fundamental

differences. The main similarities are:

1. The major mechanism in both MSE and soil nailed retaining structures is the development of tensile forces in the reinforcements due to frictional interaction and, consequently, restrainment of lateral deformations of the structures.

2. The reinforced soil mass is separated into two zones based on the points of maximum tension in the reinforcement (Fig. 8.54): an “active” zone close to the facing, where the shear stresses exerted on the surface of the reinforcement are directed outward and have a tendency to pull out the reinforcements, and a “resistant” zone, where the shear stresses are directed inward and prevent the sliding of the reinforcements.

3. The reinforcement forces are sustained by a frictional bond between the soil and the reinforcing element; the reinforced zone is stable and resists the thrust from the unreinforced soil it supports, much like a gravity retaining structure.

4. The facing of the retained structure is relatively thin, with prefabricated elements used for MSE walls and, usually, shotcrete for soil nailed walls.

The main differences are:

1. The construction procedure. Although at the end of construction the two structures may look similar, the construction sequence is radically different. Soil nailed walls are constructed “top down” by staged excavations, while MSE walls are constructed “bottom up.” Thus, the wall deformation pattern is different during construction. This also results in differences in the distribution of the forces that develop in the rein­forcement, particularly during the construction period. In an MSE structure (built bottom up), the working forces that develop in the reinforcement layers generally increase from top to bottom. In a nailed structure (built top down), the working loads that develop in the reinforcement layers are generally of uniform magnitude, similar to those in a braced excavation.

2. Nature of the soil. Soil nailing is an in situ reinforcement technique exploiting natural ground, the properties of which cannot be preselected and controlled as they are for MSE fills. MSE walls usually utilize clean, low-water-content granular backfills, which have a known friction angle and little to no cohesion. On the con­trary, nails are installed into soil and rock (natural ground) whose strength proper­ties (friction angle and cohesion) and water content can vary through a wide range.

3. Soil-reinforcement bond. Grouting techniques are usually employed to bond the nail reinforcement to the surrounding ground, with the load transferred along the grout to the soil interface. In MSE structures, friction is generated directly along the reinforcement-soil interface.

Most research to date has been done in homogeneous soils. However, there is no rea­son why the concept cannot be applied to heterogeneous soil masses if proper consid­eration of soil properties is made and rationally applied to the selection of nail length and spacing.

To be economical, soil nailed walls should be constructed in ground that can stand unsupported on a vertical or steeply sloped cut of 3 to 6 ft (1 to 2 m) for one to two days, and can maintain an open drill hole for a few hours. Soils considered favorable to soil nailing are as follows:

1. Naturally cohesive materials (silts and low-plasticity clays that are not prone to creep)

2. Naturally cemented sands and gravels

3. Weathered rock

4. Fine to medium, homogeneous sand with capillary cohesion of 60 to 100 lb/ft2 (2.9 to 4.8 kPa) associated with a water content of at least 5 to 6 percent

According to FHWA Report RD-89-108, soil nailing is generally not considered cost – effective or applicable in the following soils:

1. Loose granular soils with field standard penetration N values lower than about 10 or relative densities of less than 30 percent

2. Granular cohesionless soils of uniform size (poorly graded) with uniformity coeffi­cient (D60/D10) less than 2, unless found to be very dense; nailing of these soils may be impractical because of the necessity of stabilizing the cut face (by grouting or another permanent technique) prior to excavation

3. Soft cohesive soils with undrained shear strengths of less than 500 lb/ft2 (24 kPa), because of the inability to develop adequate pullout resistance

4. Highly plastic clays (LL > 50 and PI > 20 percent), because of their potential for excessive creep deformation

5. Expansive (swelling) and highly frost-susceptible soils

Soil nailing is also not recommended for the following conditions:

1. In ground with water pressure present at the face

2. Below the groundwater table, unless the slope can be effectively dewatered prior to excavation

3. Loose fill, granular soil with no apparent cohesion

8.8.1 Development and General Considerations

Figure 8.52 shows a cross-section of the first soil nailed wall, which was a temporary wall built in France (1972-1973) for a railroad project. Such walls are constructed from the top down during excavation. Reinforcing bars are either inserted in drilled holes and grouted into place, or driven into place. Then a facing of cast-in-place concrete or shotcrete is installed as the work progresses.

The wall in France was built in Fontainebleau sand using a high density of short nails of two different lengths: 13 ft (4 m) for nails in the upper portion of the wall and 20 ft (6 m) for those in the lower portion. The first full-size experimental wall was constructed in Germany in 1979 using grouted nails and loaded to failure. In 1981, a prefabricated – concrete-facing soil nailed wall was used in a commercial application in France. An extensive national research project conducted in France during the years 1986-1990 resulted in a noted publication titled Recommendations Clouterre.

A soil nailed wall is constructed as an integral part of the construction of an excava­tion as illustrated in Fig. 8.53. (See Art. 8.8.7.) The soil is reinforced as the slope excavation progresses. Reinforcement generally consists of bars inserted parallel to one another and placed at a downward-sloping angle. The bars are inserted in a passive state; however, as the skin friction between the soil and the nails is mobilized, the nails are placed into tension. Figure 8.54 compares the action of soil nails and ground anchors.

The work is carried out from the top downward in increments, gradually building up a reinforced soil mass. Some type of facing is generally necessary to keep the soil from caving between the soil nails. In the case of the Fontainebleau sand (effective friction angle of 38° and some cohesion), the distance for stability of the excavation

Excavate unsupported cut – 3 to 6 ft high
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STEP 4: Place initial shotcrete layer and install bearing plates and nuts

between soil nails was about 6.6 ft (2 m) with failure occurring at about 9.8 ft (3 m). The sand between the nails will slough until an “arching” action occurs within the soil. The point where this action can no longer occur because the internal friction capacity of the soil has been exceeded defines the temporary facing limit.

The concept of soil nailing has not been patented, nor is it patentable; however, numerous technologies have been patented. For soil nail walls to be cost-effective, the ground must be capable of standing unsupported while the nails and shotcrete are installed. The success of soil nail walls depends upon:

1. Selection of good applications in ground suitable for nailing

2. Ability to quickly respond to changed ground conditions

3. Use of a rational design procedure for the wall and each of its components

4. Use of good construction specifications

5. Ability of the owner and contractor to work together in a partnering concept

6. Handling of work performance on-site by knowledgeable personnel representing each of the parties, including the owner

Anchor design includes the selection of a feasible anchor system, estimation of anchor capacity, determination of unbonded length, and consideration of corrosion protection. In determining the feasibility of employing anchors at a particular location, considera­tions include the availability of underground easements, proximity of buried facilities to anchor locations, and the suitability of subsurface soil and rock conditions within the anchor stressing zone.

Ultimate anchor capacity per unit length may be estimated from Tables 8.9 and 8.10 for soil and rock, respectively. The values are based on straight-shaft anchors

installed in small-diameter holes using a low grout pressure. Other anchor types and installation procedures may result in different anchor capacities. Allowable anchor capacity for small-diameter anchors may be estimated by multiplying the ultimate anchor capacity per unit length by the bonded (or stressing) length and dividing by a factor of safety. AASHTO suggests 2.5 for anchors in soil and 3.0 for anchors in rock.

Bearing elements for anchors must be designed so that shear stresses in the vertical wall elements and facing are within allowable limits. The capacity of each anchor should be verified as part of a stressing and testing program.

Determination of the unbonded anchor length should consider the location of the critical failure surface farthest from the wall, the minimum length required to ensure minimal loss of anchor prestress due to long-term ground movements, and the depth to adequate anchoring strata. As shown in Fig. 8.49, the unbonded (or free) anchor length should not be less than 15 ft (4.6 m) and should extend 5 ft (1.5 m) or one-fifth of the design wall height, whichever is greater, beyond the critical failure surface in the soil mass being retained by the wall. For granular soils or drained cohesive soils, the critical failure surface is typically assumed to be the active failure wedge. This wedge is defined by a plane extending upward from the base of the wall at an angle of 45° + ф72 from the horizontal, where ф’ is the effective angle of soil friction. Longer free lengths may be required for anchors in plastic soils or where critical failure sur­faces are defined by planes or discontinuities with other orientations.

Selection of anchor inclination should consider the location of suitable soil or rock strata, the presence of buried utilities or other geometric constraints, and constructibility of the anchor drill holes. AASHTO suggests that anchors be located on a minimum inclination of 10° below horizontal and the bonded zone be located a minimum depth of 15 ft (4.6 m) below the ground surface. The component of vertical load resulting from anchor inclination must be included in evaluating the end bearing and settlement of vertical wall elements.

AASHTO suggests that the minimum horizontal spacing of anchors be either 3 times the diameter of the bonded zone or 4 ft (1.2 m), whichever is larger. If small spacings are required, consideration can be given to different anchor inclinations between alter­nating anchors.